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张家口市领秀城小区二期9#住宅楼,张家口市,领秀城,小区,住宅楼
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河北建筑工程学院2015-2016年第2 学期实 习 日 志系 别 土木工程学院 专 业 土木工程 班 级 工123 姓 名 王泽璐 学 号 2012304323 实习日期 2016.04.25-2106.04.29 实习周数 一周 实习地点 张家口 指导教师 庞润泽 日期实习内容20.6.04.25毕业设计进行到中期，今天跟随庞老师及同组成员共12名来到张家口一中旁边的领秀城小区工地进行参观实习。今天我们实习的单位是领秀城，这个工程位于张家口市桥西区第一中学西面，小区正在阶段性施工，建筑面积174121.7平方米，由张家口建筑设计院有限公司设计，张家口市凯博风房地产开发的118号楼，其中大多数为住宅楼，还有局部的商品用房。之后方可进行下道工序。 首先我们看到的是基础措台的处理由于室外地坪位于一个斜坡上，其中一层为商业部分，为了避免一层商业区台阶离地过高所以进行措台处理，基础高度往下递减。同时也可以看到靠近基础边上的白线区域为人工作业区，基础较浅，并且处于非雨季，直臂开挖的，边坡并未做特殊处理。日期实习内容20.6.04.26今天老师又带领我们参观的是某一楼层地基做钎探试验的现场。 地基钎探是一种土层探测施工工艺，将标志刻度的标准直径钢钎，一般采用机械方法，使用标定重量的击锤，垂直击打进入地基土层，根据钢钎进入地基土层所需的击锤数，探测土层内隐蔽构造，粗略估算土层的容许承载力。经过庞老师的讲解和工人的演示，我们了解到钎探机器的实际操作方法和钎探位置的选择，这个场地根据一定的间距，按照梅花桩的规律选择钎探位置。我们可以看到，场地上许多钎探过的位置都用标有数字标号的砖覆盖，老师说，等钎探完这些位置还会进一步处理。钎探试验的目的在于测地下是否有防空洞以及确保地下土的均匀性，没有“特殊地况”图为做钎探试验的仪器，仪器侧面有简单的人工画的尺子刻度，边于施工人员记录每当钎探仪器往下走300mm时捶击数量。做完之后将砖含数字一面翻过来，以免混淆。做完钎探试验后需通知甲方、质监站复检之后方可进行下道工序日期实习内容2016.04.27今天我们对无量楼盖的施工过程进行了参观，无梁楼盖不设梁，是一种双向受力的板柱结构，这是我第一次接触到这样的结构，这也引起了我很大的兴趣。无梁楼盖的主要特点是使同高的楼层增大净空，节省建材，提高施工进度，抗压性更高，抗振动冲击更强。需要在楼板板底布置钢丝网片，现在板模板上做白线标志处横纵向小梁的位置，间距要准确。绑扎小梁的钢筋，与主梁交接处要做好节点的搭接工作。无梁楼盖采用混凝土预制空心板，空心楼盖技术可广泛的应用于大跨度、大空间、大荷载的建筑中，具有可节省混凝土量，减轻结构的自重，降低综合造价，同时可以提高楼层的隔音效果等优点。中间预留孔洞有利于浇筑混凝土时，浇入板底的混凝土保护层更加均匀。日期实习内容2016.04.28今天主要学习参观的是框剪结构，砖混结构的施工过程。经过观察我发现该工程地上部分承重墙采用烧结页岩多孔砖，非承重墙采用加气混凝土砌块。根据老师讲解得知砌筑工程质量的基本要求是：横平竖直、厚薄均匀、砂浆饱满，上下错缝、内外搭砌、接茬可靠。其中水平灰缝过厚不仅易使砖块浮滑而且灰缝越厚则砖块的拉力越大，则砌体强度降低越多。马牙槎可分为接缝马牙槎和结构马牙槎，接缝马牙槎主要为了后续施工方便，当墙体太长时会留置。结构马牙槎是为了在马牙槎里边浇筑混凝土而成为构造柱。马牙槎的作用主要是为了提高墙体与构造柱的整体性，提高建筑物的抗震性能。在钢筋混凝土圈梁和砌体砖交接处即砌块墙顶砖采用斜砌的方法。原因一可以弥补砖砌块砌筑完成后自身变形而导致的上部出现裂缝，其次当上部梁因为受力作用而变形时这样不会使砌体墙受压而产生破坏，再者可以使墙体顶部更密实，抹灰后不易开裂。后砌的非承重的墙应该沿墙高每隔500mm配置2根6的钢筋和承重墙或柱进行拉结，每边伸入墙内不应小于500mm。钢筋砼结构中的砌体填充墙，应与柱墙脱开或采用柔性连接，并应该符合下列的要求：1） 填充墙在平面与竖向布置，宜均匀对称，以避免形成薄弱层或短柱。2） 砌体的砂浆强度等级不应该低于M5，墙顶应沿框架柱全高每隔500mm设26拉筋，拉筋伸入墙内的长度，6、7度时不应小于墙长的1/5且不小于700mm，8、9度时宜沿墙全长贯通；虽然未规定拉结筋必须在端部设置弯钩，但在结构设计说明中一般都画出了180的弯钩。 日期实习内容2016.04.29今天是实习的最后一天，我对这几天没有提到的实习的经历进行了整理，下图是工人被工人截断的桩头，这些桩为干塑混凝土挤密桩，通过膨胀挤压作用，使整个地基土密实性更好，从而降低黄土的湿陷性以及提高地基承载力。本工程采用的钢筋连接方式有机械连接、绑扎搭接及焊接的方式，主要采用的有绑扎搭接。老师又着重强调的剪力墙水平筋和垂直筋的布置位置。还有板钢筋跨度短的在下面，跨度长的布置在上面，挡土墙钢筋外侧横向钢筋在墙竖向钢筋内侧，在内墙外侧横向钢筋在竖向钢筋外侧，内侧横向钢筋在墙外侧，水池的墙外侧横向钢筋均在竖向钢筋内侧，同样横向内侧钢筋在竖向钢筋内侧。 梁下部钢筋的绑扎，混凝土的浇筑应该成拱形或者水平，此图中梁中部已经下弯，为防止这种情况在绑筋时应该留出弧度，浇筑前在下部搭设脚手架并做好支撑，防止梁板变形。 今天是参观实习的最后一天了，实习虽然很累，但是通过这几天的参观实习，我了解到了许多施工现场的知识，从挖土方到主体的建筑及二次施工都见识到了许多。这对我以后更好的工作学习都有非常重要的作用，最后也感谢老师的辛勤教导!翻 译河北建筑工程学院毕业设计（论文）外文资料翻译学院： 土木工程学院 专业： 土木工程 班级： 工123 姓名： 王泽璐 学号： 2012304323 指导老师： 庞润泽 外文出处：The Engineering Index附 件：1、外文原文；2、外文资料翻译译文。指导教师评语： 签字： 年 月 日 Cyclic behavior of steel moment frameconnections under varying axial load and lateral displacementsAbstract: This paper discusses the cyclic behavior of four steel moment connections tested under variable axial load and lateral displacements. The beam specim- ens consisted of a reducedbeam section, wing plates and longitudinal stiffeners. The test specimens were subjected to varying axial forces and lateral displace- ments to simulate the effects on beams in a Coupled-Girder Moment-Resisting Framing system under lateral loading. The test results showed that the specim-ens responded in a ductile manner since the plastic rotations exceeded 0.03 rad without significant drop in the lateral capacity. The presence of the longitudin-al stiffener assisted in transferring the axial forces and delayed the formation of web local buckling.1. Introduction Aimed at evaluating the structural performance of reduced-beam section(RBS) connections under alternated axial loading and lateral displacement, four full-scale specimens were tested. These tests were intended to assess the performance of the moment connection design for the Moscone Center Exp- ansion under the Design Basis Earthquake (DBE) and the Maximum Considered Earthquake (MCE). Previous research conducted on RBS moment connections 1,2 showed that connections with RBS profiles can achieve rotations in excess of 0.03 rad. However, doubts have been cast on the quality of the seismic performance of these connections under combined axial and lateral loading. The Moscone Center Expansion is a three-story, 71,814 m2 (773,000 ft2) structure with steel moment frames as its primary lateral force-resisting system. A three dimensional perspective illustration is shown in Fig. 1. The overall height of the building, at the highest point of the exhibition roof, is approxima- tely 35.36 m (116ft) above ground level. The ceiling height at the exhibition hall is 8.23 m (27 ft) , and the typical floor-to-floor height in the building is 11.43 m (37.5 ft). The building was designed as type I according to the requi- rements of the 1997 Uniform Building Code. The framing system consists of four moment frames in the EastWest direct- ion, one on either side of the stair towers, and four frames in the NorthSouth direction, one on either side of the stair and elevator cores in the east end and two at the west end of the structure 4. Because of the story height, the con- cept of the Coupled-Girder Moment-Resisting Framing System (CGMRFS) was utilized. By coupling the girders, the lateral load-resisting behavior of the moment framing system changes to one where structural overturning moments are resisted partially by an axial compressiontension couple across the girder system, rather than only by the individual flexural action of the girders. As a result, a stiffer lateral load resisting system is achieved. The vertical element that connects the girders is referred to as a coupling link. Coupling links are analogous to and serve the same structural role as link beams in eccentrically braced frames. Coupling links are generally quite short, having a large shear- to-moment ratio. Under earthquake-type loading, the CGMRFS subjects its girders to wariab- ble axial forces in addition to their end moments. The axial forces in the Fig. 1. Moscone Center Expansion Project in San Francisco, CAgirders result from the accumulated shear in the link. Fig 2.Analytical model of CGMRF Nonlinear static pushover analysis was conducted on a typical one-bay model of the CGMRF. Fig. 2 shows the dimensions and the various sections of the model. The link flange plates were 28.5 mm 254 mm (1 1/8 in 10 in) and the web plate was 9.5 mm 476 mm (3 /8 in 18 3/4 in). The SAP 2000 computer program was utilized in the pushover analysis 5. The frame was characterized as fully restrained(FR). FR moment frames are those frames for 1170which no more than 5% of the lateral deflections arise from connection deformation 6. The 5% value refers only to deflection due to beamcolumn deformation and not to frame deflections that result from column panel zone deformation 6, 9. The analysis was performed using an expected value of the yield stress and ultimate strength. These values were equal to 372 MPa (54 ksi) and 518 MPa (75 ksi), respectively. The plastic hinges loaddeformation behavior was approximated by the generalized curve suggested by NEHRP Guidelines for the Seismic Rehabilitation of Buildings 6 as shown in.Fig. 3. y was calcu- lated based on Eqs. (5.1) and (5.2) from 6, as follows:PM hinge loaddeformation model points C, D and E are based on Table 5.4 from 6 for y was taken as 0.01 rad per Note 3 in 6, Table 5.8. Shear hinge load- loaddeformation model points C, D and E are based on Table 5.8 6, Link Beam, Item a. A strain hardening slope between points B and C of 3% of the elastic slope was assumed for both models. The following relationship was used to account for momentaxial load interaction 6:where MCE is the expected moment strength, ZRBS is the RBS plastic section modulus (in3), is the expected yield strength of the material (ksi), P is the axial force in the girder (kips) and is the expected axial yield force of the RBS, equal to (kips). The ultimate flexural capacities of the beam and the link of the model are shown in Table Fig. 4 shows qualitatively the distribution of the bending moment, shear force, and axial force in the CGMRF under lateral load. The shear and axial force in the beams are less significant to the response of the beams as compared with the bending moment, although they must be considered in design. The qualita- tive distribution of internal forces illustrated in Fig. 5 is fundamentally the same for both elastic and inelastic ranges of behavior. The specific values of the internal forces will change as elements of the frame yield and internal for- ces are redistributed. The basic patterns illustrated in Fig. 5, however, remain the same.Inelastic static pushover analysis was carried out by applying monotonicallyincreasing lateral displacements, at the top of both columns, as shown in Fig. 6. After the four RBS have yielded simultaneously, a uniform yielding in the web and at the ends of the flanges of the vertical link will form. This is the yield mechanism for the frame , with plastic hinges also forming at the base of the columns if they are fixed. The base shear versus drift angle of the model is shown in Fig. 7 . The sequence of inelastic activity in the frame is shown on the figure. An elastic component, a long transition (consequence of the beam plastic hinges being formed simultaneously) and a narrow yield plateau characterize the pushover curve. The plastic rotation capacity, qp, is defined as the total plastic rotation beyond which the connection strength starts to degrade below 80% 7. This definition is different from that outlined in Section 9 (Appendix S) of the AISC Seismic Provisions 8, 10. Using Eq. (2) derived by Uang and Fan 7, an estimate of the RBS plastic rotation capacity was found to be 0.037 rad:Fyf was substituted for RyFy 8, where Ry is used to account for the differ- ence between the nominal and the expected yield strengths (Grade 50 steel, Fy=345 MPa and Ry =1.1 are used).3. Experimental program The experimental set-up for studying the behavior of a connection was based on Fig. 6(a). Using the plastic displacement dp, plastic rotation gp, and plastic story drift angle qp shown in the figure, from geometry, it follows that:And: in which d and g include the elastic components. Approximations as above are used for large inelastic beam deformations. The diagram in Fig. 6(a) suggest that a sub assemblage with displacements controlled in the manner shown in Fig. 6(b) can represent the inelastic behavior of a typical beam in a CGMRF. The test set-up shown in Fig. 8 was constructed to develop the mechanism shown in Fig. 6(a) and (b). The axial actuators were attached to three 2438 mm 1219 mm 1219 mm (8 ft 4 ft 4 ft) RC blocks. These blocks were tensioned to the laboratory floor by means of twenty-four 32 mm diameter dywidag rods. This arrangement permitted replacement of the specimen after each test. Therefore, the force applied by the axial actuator, P, can be resolved into two or thogonal components, Paxial and Plateral. Since the inclination angle of the axial actuator does not exceed 3.0, therefore Paxial is approximately equal to P 4. However, the lateral component, Plateral, causes an additional moment at the beam-to column joint. If the axial actuators compress the specimen, then the lateral components will be adding to the lateral actuator forces, while if the axial actuators pull the specimen, the Plateral will be an opposing force to the lateral actuators. When the axial actuators undergoaxial actuators undergo a lateral displacement _, they cause an additional moment at the beam-to-column joint (P- effect). Therefore, the moment at the beam-to column joint is equal to: where H is the lateral forces, L is the arm, P is the axial force and _ is the lateral displacement. Four full-scale experiments of beam column connections were conducted. The member sizes and the results of tensile coupon tests are listed in Table 2All of the columns and beams were of A572 Grade 50 steel (Fy 344.5 MPa). The actual measured beam flange yield stress value was equal to 372 MPa (54 ksi), while the ultimate strength ranged from 502 MPa (72.8 ksi) to 543 MPa (78.7 ksi). Table 3 shows the values of the plastic moment for each specimen (based on measured tensile coupon data) at the full cross-section and at the reduced section at mid-length of the RBS cutout. The specimens were designated as specimen 1 through specimen 4. Test specimens details are shown in Fig. 9 through Fig. 12. The following features were utilized in the design of the beamcolumn connection: The use of RBS in beam flanges. A circular cutout was provided, as illustr- ated in Figs. 11 and 12. For all specimens, 30% of the beam flange width was removed. The cuts were made carefully, and then ground smooth in a direct- tion parallel to the beam flange to minimize notches. Use of a fully welded web connection. The connection between the beam web and the column flange was made with a complete joint penetration groove weld (CJP). All CJP welds were performed according to AWS D1.1 Structural Welding CodeUse of two side plates welded with CJP to exterior sides of top and bottom beam flan- ges, from the face of the column flange to the beginning of the RBS, as shown in Figs. 11 and 12. The end of the side plate was smoothed to meet the beginning of the RBS. The side plates were welded with CJP with the column flanges. The side plate was added to increase the flexural capacity at the joint location, while the smooth transition was to reduce the stress raisers, which may initiate fractureTwo longitudinal stiffeners, 95 mm 35 mm (3 3/4 in 1 3/8 in), were welded with 12.7 mm (1/2 in) fillet weld at the middle height of the web as shown in Figs. 9 and 10. The stiffeners were welded with CJP to column flanges. Removal of weld tabs at both the top and bottom beam flange groove welds. The weld tabs were removed to eliminate any potential notches introduced by the tabs or by weld discontinuities in the groove weld run out regions. Use of continuity plates with a thickness approximately equal to the beam flange thickness. One-inch thick continuity plates were used for all specimens.While the RBS is the most distinguishing feature of these test specimens, the longitudinal stiffener played an important role in delaying the formation of web local buckling and developing reliable connection performance4. Loading history Specimens were tested by applying cycles of alternated load with tip displacement increments of _y as shown in Table 4. The tip displacement of the beam was imposed by servo-controlled actuators 3 and 4. When the axial force was to be applied, actuators 1 and 2 were activated such that its force simulates the shear force in the link to be transferred to the beam. The variable axial force was increased up to 2800 kN (630 kip) at +0.5_y. After that, this lo- ad was maintained constant through the maximum lateral displacement. maximum lateral displacement. As the specimen was pushed back the axial force remained constant until 0.5 y and then started to decrease to zero as the specimen passed through the neutral position 4. According to the upper bound for beam axial force as discussed in Section 2 of this paper, it was concluded that P =2800 kN (630 kip) is appropriate to investigate this case in RBS loading. The tests were continued until failure of the specimen, or until limitations of the test set-up were reached. 5. Test results The hysteretic response of each specimen is shown in Fig. 13 and Fig. 16. These plots show beam moment versus plastic rotation. The beam moment is measured at the middle of the RBS, and was computed by taking an equiva- lent beam-tip force multiplied by the distance between the centerline of the lateral actuator to the middle of the RBS (1792 mm for specimens 1 and 2, 3972 mm for specimens 3 and 4). The equivalent lateral force accounts for the additional moment due to P effect. The rotation angle was defined as the lateral displacement of the actuator divided by the length between the centerline of the lateral actuator to the mid length of the RBS. The plastic rotation was computed as follows 4:where V is the shear force, Ke is the ratio of V/q in the elastic range. Measurements and observations made during the tests indicated that all of the plastic rotation in specimen 1 to specimen 4 was developed within the beam. The connection panel zone and the column remained elastic as intended by design. 5.1. Specimens 1 and 2 The responses of specimens 1 and 2 are shown in Fig. 13. Initial yielding occurred during cycles 7 and 8 at 1_y with yielding observed in the bottom flange. For all test specimens, initial yielding was observed at this location and attributed to the moment at the base of the specimen 4. Progressing through the loading history, yielding started to propagate along the RBS bottom flange. During cycle 3.5_y initiation of web buckling was noted adjacent to the yielded bottom flange. Yielding started to propagate along the top flange of the RBS and some minor yielding along the middle stiffener. During the cycle of 5_y with the increased axial compression load to 3115 KN (700 kips) a severe web buckle developed along with flange local buckling. The flange and the web local buckling became more pronounced with each successive loading cycle. It should be noted here that the bottom flange and web local buckling was not accompanied by a significant deterioration in the hysteresis loops. A crack developed in specimen 1 bottom flange at the end of the RBS where it meets the side plate during the cycle 5.75_y. Upon progressing through the loading history, 7_y, the crack spread rapidly across the entire width of the bottom flange. Once the bottom flange was completely fractured, the web began to fracture. This fracture appeared to initiate at the end of the RBS，then propagated through the web net section of the shear tab, through the middle stiffener and the through the web net section on the other side of the stiffener. The maximum bending moment achieved on specimen 1 during theDuring the cycle 6.5 y, specimen 2 also showed a crack in the bottom flange at the end of the RBS where it meets the wing plate. Upon progressing thou- gh the loading history, 15 y, the crack spread slowly across the bottom flan- ge. Specimen 2 test was stopped at this point because the limitation of the test set-up was reached.The maximum force applied to specimens 1 and 2 was 890 kN (200 kip). The kink that is seen in the positive quadrant is due to the application of the varying axial tension force. The load-carrying capacity in this zone did not deteriorate as evidenced with the positive slope of the forcedisplacement curve. However, the load-carrying capacity deteriorated slightly in the neg- ative zone due to the web and the flange local buckling. Photographs of specimen 1 during the test are shown in Figs. 14 and 15. Severe local buckling occurred in the bottom flange and portion of the web next to the bottom flange as shown in Fig. 14. The length of this buckle extended over the entire length of the RBS. Plastic hinges developed in the RBS with extensive yielding occurring in the beam flanges as well as the web. Fig. 15 shows the crack that initiated along the transition of the RBS to the side wing plate. Ultimate fracture of specimen 1 was caused by a fracture in the bottom flange. This fracture resulted in almost total loss of the beam- carrying capacity. Specimen 1 developed 0.05 rad of plastic rotation and showed no sign of distress at the face of the column as shown in Fig. 15. 5.2. Specimens 3 and 4 The response of specimens 3 and 4 is shown in Fig. 16. Initial yielding occured during cycles 7 and 8 at 1_y with significant yielding observed in the bottom flange. Progressing through the loading history, yielding started to propagate along the bottom flange on the RBS. During cycle 1.5_y initiation of web buckling was noted adjacent to the yielded bottom flange. Yielding started to propagate along the top flange of the RBS and some minor yielding along the middle stiffener. During the cycle of 3.5_y a severe web buckle developed along with flange local buckling. The flange and the web local buckling bec- ame more pronounced with each successive loading cycle. During the cycle 4.5 y, the axial load was increased to 3115 KN (700 kips) causing yielding to propagate to middle transverse stiffener. Progressing through the loading history, the flange and the web local buckling became more severe. For both specimens, testing was stopped at this point due to limitations in the test set-up. No failures occurred in specimens 3 and 4. However, upon removing specimen 3 to outside the laboratory a hairline crack was observed at the CJP weld of the bottom flange to the column. The maximum forces applied to specimens 3 and 4 were 890 kN (200 kip) and 912 kN (205 kip). The load-carrying capacity deteriorated by 20% at the end of the tests for negative cycles due to the web and the flange local buckling. This gradual reduction started after about 0.015 to 0.02 rad of plastic rotation. The load-carrying capacity during positive cycles (axial tension applied in the girder) did not deteriorate as evidenced with the slope of the forcedisplacement envelope for specimen 3 shown in Fig. 17. A photograph of specimen 3 before testing is shown in Fig. 18. Fig. 19 is a Fig. 16. Hysteretic behavior of specimens 3 and 4 in terms of moment at middle RBS versus beam plastic rotation.photograph of specimen 4 taken after the application of 0.014 rad displacem- ent cycles, showing yielding and local buckling at the hinge region. The beam web yielded over its full depth. The most intense yielding was observed in the web bottom portion, between the bottom flange and the middle stiffener. The web top portion also showed yielding, although less severe than within the bottom portion. Yielding was observed in the longitudinal stiffener. No yiel- ding was observed in the web of the column in the joint panel zone. The un- reduced portion of the beam flanges near the face of the column did not show yielding either. The maximum displacement applied was 174 mm, and the maximum moment at the middle of the RBS was 1.51 times the plastic mom ent capacity of the beam. The plastic hinge rotation reached was about 0.032 rad (the hinge is located at a distance 0.54d from the column surface,where d is the depth of the beam). 5.2.1. Strain distribution around connection The strain distribution across the flangesouter surface of specimen 3 is shown in Figs. 20 and 21. The readings and the distributions of the strains in specimens 1, 2 and 4 (not presented) showed a similar trend. Also the seque- nce of yielding in these specimens is similar to specimen 3.The strain at 51 mm from the column in the top flangeouter surface remained below 0.2% during negative cycles. The top flange, at the same location, yielded in compression only.The longitudinal strains along the centerline of the bottomflange outer face are shown in Figs. 22 and 23 for positive and negative cycles, respectively. From Fig.23, it is found that the strain on the RBS becomes several times larg- er than that near the column after cycles at 1.5_y; this is responsible for the flange local buckling. Bottom flange local buckling occurred when the average strain in the plate reached the strain-hardening value (esh _ 0.018) and the reduced-beam portion of the plate was fully yielded under longitudinal stresses and permitted the development of a full buckled wave. 5.2.2. Cumulative energy dissipated The cumulative energy dissipated by the specimens is shown in Fig. 24. The cumulative energy dissipated was calculated as the sum of the areas enclosed the lateral loadlateral displacement hysteresis loops. Energy dissipation sta- rted to increase after cycle 12 at 2.5 y (Fig. 19). At large drift levels, energy dissipation augments significantly with small changes in drift. Specimen 2 dissipated more energy than specimen 1, which fractured at RBS transition. However, for both specimens the trend is similar up to cycles at q =0.04 radIn general, the dissipated energy during negative cycles was 1.55 times bigger than that for positive cycles in specimens 1 and 2. For specimens 3 and 4 the dissipated energy during negative cycles was 120%, on the average, that of the positive cycles.The combined phenomena of yielding, strain hardening, in-plane and out- of-plane deformations, and local distortion all occurred soon after the bottom flange RBS yielded. 6. Conclusions Based on the observations made during the tests, and on the analysis of the instrumentation, the following conclusions were developed:1. The plastic rotation exceeded the 3% radians in all test specimens.2. Plastification of RBS developed in a stable manner.3. The overstrength ratios for the flexural strength of the test specimens were equal to 1.56 for specimen 1 and 1.51 for specimen 4. The flexural strength capacity was based on the nominal yield strength and on the FEMA-273 beamcolumn equation.4. The plastic local buckling of the bottom flange and the web was not accompanied by a significant deterioration in the load-carrying capacity.5. Although flange local buckling did not cause an immediate degradation ofstrength, it did induce web local buckling.6. The longitudinal stiffener added in the middle of the beam web assisted in transferring the axial forces and in delaying the formation of web local buckling. How ever, this has caused a much higher overstrength ratio, which had a significant impact on the capacity design of the welded joints, panel zone and the column.7. A gradual strength reduction occurred after 0.015 to 0.02 rad of plastic rotation during negative cycles. No strength degradation was observed during positive cycles.8. Compression axial load under 0.0325Py does not affect substantially the connection deformation capacity.9. CGMRFS with properly designed and detailed RBS connections is a reliable system to resist earthquakes.出自工程索引，The Engineering Index，简称EI。弯钢框架结点在变化轴向荷载和侧向位移的作用下的周期性行为摘要:这篇论文讨论的是在变化的轴向荷载和侧向位移的作用下，接受测试的四种受弯钢结点的周期性行为。梁的试样由变截面梁，翼缘以及纵向的加劲肋组成。受测试样加载轴向荷载和侧向位移用以模拟侧向荷载对组合梁抗弯系统的影响。实验结果表明试样在旋转角度超过0.03弧度后经历了从塑性到延性的变化。纵向加劲肋的存在帮助传递轴向荷载以及延缓腹板的局部弯曲。1、引言 为了评价变截面梁（RBS）结点在轴向荷载和侧向位移下的结构性能，对四个全尺寸的样品进行了测试。这些测试打算评价为旧金山展览中心扩建设计的受弯结点在满足设计基本地震等级（DBE）和最大可能地震等级（MCE）下的性能。基于上述而做的对RBS受弯结点的研究指出RBS形式的结点能够获得超过0.03弧度的旋转角度。然而，有人对于这些结点在轴向和侧向荷载作用下的抗震性能质量提出了怀疑。 旧金山展览中心扩建工程是一个3层构造，并以钢受弯框架作为基本的侧向力抵抗系统。Fig.1是一幅三维透视图。建筑的总标高为展览厅屋顶的最高点，大致是35.36m（116ft）。展览厅天花板的高度是8.23m（27ft），层高为11.43m（37.5ft）。建筑物按照1997统一建筑规范设计。框架系统由以下几部分组成：四个东西走向的受弯框架，每个电梯塔边各一个；四个南北走向的受弯框架，在每个楼梯和电梯井各一个的；整体分布在建筑物的东西两侧。考虑到层高的影响，提出了双梁抗弯框架系统的观念。通过连接大梁， 受弯框架系统的抵抗荷载的行为转化为结构倾覆力矩部分地被梁系统的轴向压缩-拉伸分担，而不是仅仅通过梁的弯曲。结果，达到了一个刚性侧向荷载抵抗系统。竖向部分与梁以联结杆的形式连接。联结杆在结构中模拟偏心刚性构架并起到与其相同的作用。通常地联结杆都很短，并有很大的剪弯比。 在地震类荷载的作用下，CGMRFS梁的最终弯矩将考虑到可变轴向力的影响。梁中的轴向力是切向力连续积累的结果。2CGMRF的解析模型 非线性静力推出器模型是以典型的单间CGMRF模板为指导。图2展示了模型的尺寸规格和多个部分。翼缘板尺寸为28.5mm 254mm（1 1/8in 10in），腹板尺寸为9.5mm 476mm（3/8in 18 3/4in）。推进器模型中运用了SAP 2000计算机程序。框架的特色是全约束（FR）。FR受弯框架是一种由结点应变引起的挠度不超过侧向挠度的5%的框架。这个5%仅与梁-柱应变有关，而与柱底板区应变引起的框架应变无关。模型通过屈服应力和匹配强度的期望值来运行。这些值各自为372Mpa（54ksi）和518Mpa（75ksi）。Fig.3显示了塑性铰的荷载-应变行为是通过建筑物地震恢复的NEHRP指标以广义曲线的形式逼近的。 y以Eps5.1和5.2为基底运算，如下： P-M铰合线荷载-应变模型上的点C，D和E的取值如表5.4y以0.01rad为幅度取值见表5.8。切变铰合线荷载-应变模型点C，D和E取值见表5.8。对于连续梁，假定两个模型点B和C之间的形变硬化比有3%的弹性比。Fig.4定性的给出了侧向荷载下的CGMRF中的弯矩，切应力和正应力的分布。其中切应力和正应力对梁的影响要小于弯矩的作用，尽管他们必须在设计中加以考虑。内力分布图解见Fig.5，可见，弹性范围和非弹性范围的内力行为基本相同。内力的比值将随框架的屈服和内力的重分布的变化而变化。基本内力图见Fig.5，然而，仍然是一样的。 非静力推进器模型的运行通过柱子顶部的侧向位移的单调增加来实现，如Fig.5所示。在四个RBS同时屈服后，发生在腹板与翼缘端部的竖向的统一屈服将开始形成。这是框架的屈服中心，在柱子被固定后将在柱底部形成塑性铰。Fig.7给出了基本切应力偏移角。图中还给出了框架中非弹性活动的次序。对于一个弹性组成，推进器将有一个特有的很长的过渡（同时形成塑性铰）和一个很短的屈服平稳阶段。塑性旋转能力， 被定义为：结点强度从开始递减到低于80%的总的塑性旋转角。这个定义不同于第9段（附录）AISC地震条款的描述。使用Eq源于RBS塑性旋转能力被定在0.037弧度。被 替代， 用来计算理论屈服强度与实际屈服强度的区别（标号是50钢）。3实践规划如图6所示，实验布置是为了研究基于典型的CGMRF结构下的结点在动力学中的能量耗散。用图中所给的塑性位移，塑性转角，塑性偏移角，由几何结构，有如下：这里的和包括了弹性组合。上述近似值用于大型的非弹性梁的变形破坏。图6a表明用图6b所示的位移控制下的替代组合能够表示CGMRF结构中的典型梁的非弹性行为。图8所示，建立这个实验装置来发展图6a和图6b所示的机构学。轴心装置附以3个2438mm1219mm1219mm（8ft4ft4ft）RC块。并用24个32mm径的杆与实验室的地板固定。这种装置允许在每次测验后换实验样品。根据实验布置的动力学要求，随着侧面的元件放置，轴向的元件，元件1和元件2，将钉到B和C中去， 如图8所示。因此，轴向元件提供的轴向力P可以被分解为相互正交的力的组合， 和 ，由于轴向力的倾斜角度不超过 ，因此 近似等于P。然而，侧向力分量， ，引起了一个在梁柱交接处的附加弯矩。如果轴向元件压试样的话，那么将会加到侧向力中，若轴向是拉力，对于侧向元件来说则是个反向力。当轴向元件有个侧向位移 ，他们将在梁柱交接处引起一个附加弯矩，因此，梁柱交接处的弯矩等于： M=HL+P 其中H是侧向力，L是力臂，P是轴向力， 是侧向位移。 四个梁柱结点全尺寸实验做完了。拉伸试样检测的结果和构件尺寸见表2。所有柱和梁的钢筋为A572标号50钢（ =344.5Mpa）。经测定的梁翼缘屈服应力值等于372Mpa（54ksi），整体的强度范围是从502Mpa（72.8ksi）到543Mpa（78.7ksi）。表3列出了各个试样的全截面和RBS中间变截面处的塑性弯矩值（受拉应力下的数据）。 本文所指的试样专指试样1到4。被检试样细部图见图9到图12。在设计梁柱结点时用到了以下数据：梁翼缘部分采用RBS结构。配备环形掏槽，如图11和图12所示。对于所有的试样，切除30%翼缘宽度。切除工作做的十分精细，并打磨光滑且与梁翼缘保持平行以尽量见效切口。应用全焊接腹板结点。梁腹板与柱翼缘之间的结点采用全焊缝焊接（CJP）。所有CJP焊接严格依照AWS D1.1结构焊接规范。采用双侧板加CJP形式连接梁翼缘的顶部和底部和柱表面到变截面开始处，如图11和图12。侧板尾部打磨光滑以便同RBS连接。侧板采用CJP形式与柱边缘相连接。侧板的作用是增加受弯单元的承受能力，平稳过渡是为了减少应力集中而导致的破裂。 两根纵向的加劲肋，95mm35mm（3 3/4in 1 3/8 in），以12.7mm的角焊缝焊接到腹板的中间高度，如图9和10。加劲肋采用CJP的形式焊接到柱的边缘。切除梁翼缘顶部和低部的坡口焊缝处的多余焊接部分。以便消除坡口焊接断口处可能产生的断裂。除去翼缘低部的衬垫板条。以便消除衬垫板条带来的断口效应并增加安全性。使用与梁翼缘厚度近似相同的连续板。所有试样板厚均为一英寸。由于RBS是受检试样最容易区分的特征，纵向的加劲肋在延缓局部弯曲和提高结点可靠性方面扮演着重要的角色。4荷载历史 试样被加以周期性交替的荷载，其末端的位移y的增加如图4所示。梁的末端位移受伺服控制装置
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