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Tests on a Half-Scale Two-Story Seismic-Resisting Precast Concrete BuildingThis paper describes experimental studies on the seismic behavior and design of precast concrete buildings. A half-scale two-story precast concrete building incorporating a dual system and representing a parking structure in Mexico City was investigated. The structure was tested up to failure in a laboratory under simulated seismic loading. In some of the beam-to-column joints, the bottom longitudinal bars of the beam were purposely undeveloped due to dimensional constraints.Emphasis is given in the study on the evaluation of the observed global behavior of the test structure. This behavior showed that the walls of the test structure controlled the force path mechanism and significantly reduced the lateral deformation demands in the precast frames. Seismic design criteria and code implications for precast concrete structures resulting from this research are discussed. The end result of this research is that a better understanding of the structural behavior of this type of building has been gained results of simulated seismic load tests of a two story precast concrete building constructed with precast concrete elements that are used in Mexico are described herein. The structural system chosen in the test structure is the so called dual type, defined as the combination of structural walls and beam-to-column frames. Connections between precast beams and columns in the test structure are of the window type. This type of construction is typically used in low- and medium rise buildings in which columns are connected with windows at each story level. These windows contain the top and bottom reinforcement. Fig. 1 shows this type of construction for a commercial building in Mexico City. In most precast concrete frames such as those shown in Fig, 1, longitudinal beam bottom bars are not fully developed due to constraints imposed by the dimensions of file columns in beam-to-column joints. In an effort to overcome this deficiency, and as described later, some practicing engineers in Mexico design these joints by providing hoops around the hooks of that reinforcement in order to achieve its required continuity. However, this practice is not covered in the ACI Building Code (ACI318-02), nor in the Mexico City Building Code (MCBC, 1993). Part of this research was done to address this issue. The objectives of this research were Io evaluate the observed behavior of a precast concrete structures in the laboratory and to propose the use of precast structural elements or precast structures with both an acceptable level of expected seismic performance and appealing features from the viewpoint of construction Emphasis is given in this paper on the global behavior of the test structure. In the second part of this research which gill be presented in a companion paper, the observed behavior of connections between precast elements in the test structure, as well as the behavior of the precast floor system will be discussed in detail. Structural and non structural damages observed in buildings during past earthquakes throughout the world have shown the importance of controlling lateral displacement in structures to reduce building damage during earth- quakes. It is also relevant to mention that there are several cases of structures in moderate earthquakes in which the observed damage in non-structural elements in buildings was considerable even though the structural elements showed little or no damage. This behavior is also related Io excessive lateral displacement demands in the structure. To minimize seismic damage during earthquakes, the above discussion suggests the convenience of using a structural system capable of controlling lateral displacements in structures. A solution of this type is the so-called dual system. Studies by Paulay and Priestley4 on the seismic response of dual systems have shown that the presence of walls reduce the dynamic moment demands in structural elements in the frame subsystem. Also in conjunction with shake table tests conducted on a cast-in-place reinforced concrete dual system. Bertero5 has shown the potential of the dual system, in achieving excellent seismic behavior n this investigation, the dual system is applied to the case of precast concrete structures.DUCTILITY DEMAND IN DUAL SYSTEMSIn order to develop a base for a later analysis of the observed seismic response of the test structure studied in this project a simple analytical model is used to evaluate the main features of ductility demands in dual systems. Fig 2 shows the results of a simple approach to analyze the lateral load response iii a dual system. The lateral load has been normalized in such a manner that the combination of maximum lateral resistance in both subsystern i.e. walls and frames-leads to a lateral resistance of the global system equal to unity b is also assumed that both subsystems have the same maximum lateral resistance. In the first case (Fig 2a), it is assumed that the wall and frame subsystems have global displacement ductility capacities equal to 4 and 2 respectively. In the second case (Fig. 2b), the frame subsystem response is assumed to be elastic, and the lateral stiffness of the wall subsystem is taken to be 4 times that of the frame subsystem.As shown in Fig 2, the lateral deformation compatibility of the combined system is controlled by the lateral deformation capacity of the wall subsystem. In the first case Fig 2ak an elastic-plastic envelope for the lateral global response of the dual system is assumed, and the corresponding displacement ductility (u) is equal to 33.For the second case (Fig. 2b) with an elastic behavior of the frame subsystem, this ductility is equal to 25. These simple examples illustrate that in the analyzed cases, due to the higher flexibility in the frame subsystems as compared to those of the wall subsystern, in a dual system, the ductility demands in the frame subsystem result in smaller ductility values than those of the wall subsystem. This analytical finding was verified in this study from the experimental studies conducted on the test structure. This verification is later discussed in the paper It is of interest to note that results of the type shown in Fig. 2 have been also found by Bertero in shake table tests of a dual system. DESCRIPTION OF TEST STRUCTUREThe test structure used in this investigation is a two-story precast concrete building, representative of a low-rise parking structure located in the highest seismic zone of Mexico City. The prototype was constructed at one-half scale. For the sake of simplicity, ramps required in a parking structure have not been considered in the selected prototype structure. Their use, requiring large openings in the floor system, would have required a very complex model of the floor system for both linear and nonlinear analysis of the structure.A detailed description of the dimensions, materials, design procedures, and construction of the test structure can be found elsewhere.6 A summary of this information is given below. The dimensions and some characteristics of the test structure are shown in Fig. 3. The longitudinal and transverse are shown in Fig3a. Also, the exterior (longitudinal) frame containing the wall (Column Lines 1 and 3) are termed the lateral frame (see Fig, 3b), and the internal (longitudinal) frame with the single tee (Column Line 2) are termed the central frame. Doable tees spanning in the longitudinal direction are supported by L-shaped precast beams in the transverse direction as shown in Fig3a. The structure uses precast frames and precast structural walls, the latter elements functioning as the main lateral load resisting system. Fig. 4 shows an early phase of the construction of the test structure. As can be seen, the windows in the columns and walls are left in these elements for a later assemblage with the precast beams.The unfastened design base shear required by the Mexico City Building Code (MCBC, 1993)2 is 0.2WT, where WT is the total weight of the prototype structure, assuming a dead load of 5,15 KPa (108 psi) and a live load of 0.98 KPa (20.5 psi). The prototype structure was designed using procedures of elastic analyses and proportioning requirements of the MCBC, In these analyses, the gross moment of inertia of the members in the structure was considered and rigid offsets (distances from the joints to the face of the supports) were assumed for all beams in the structure except for beams in the central frame, which had substandard detailing as will be described latch. Results from these analyses indicated that the structural walls in the test structure would take about 65 percent of the design lateral loads. A review of the nominal lateral resistance of the structure using the MCBC procedures showed that this resisting force was about 1.3 times the required code lateral resistance (0,2Wr), This is one of several factors, later discussed, that contributed to the over-strength of the structure.The longitudinal reinforcement in all the structural elements of the test structore was deformed bars from Grade 420 steel. Table 1 lists the concrete compressive cylinder strengths for different members of the prototype structure. Fig. 5 shows typical reinforcing details for precast beams spanning in the direction of the applied lateral load (see Fig. 3). Figs. 6 and 7 show reinforcing details for the columns, and for the structural wails and their foundation, respectively. It should be mentioned that the test structure was designed with the requirements for moderately ductile structures specified by the MCBC. According to these provisions, the test structure did not require special structural walls with boundary elements such as those specified in Chapter 21 of AC1 318 02.The precast two-story columns were connected to the precast foundation by unthreading them in a grouted socket type connection. The reinforcing details of the foundation, as well as its design procedure and behavior in the test structure are discussed in the companion paper? Tae beam-to-cadmium joints in file test structure were cast-in-place to enable positioning the longitudinal reinforcement of the framing beams. The beam top reinforcement was placed in sum on top of the precast beams. Fig. 8 shows typical reinforcing details for the joints in the double tees of the central frame. Since these tees and their supporting L-shaped beams in Axes A or C (see Fig. 3) had the same depth, the hooked bottom longitudinal bars in the double tees could not pass through the full depth of the column because of interference with the bottom bars from file transverse beam (see Fig. g).As a result, these hooked bars possessed only about 55 percent of the development length required by Chapter 21 of ACI 318-02. In an attempt to anchor these hooked bars, some designers in Mexico provide closed hoops around the hooks, as shown in Fig. 8. The effectiveness of this approach was also studied in the companion paper.3 Beam to-column joints in the lateral frames of the test structure had transverse beams that were deeper than the longitudinal beams. This made it possible for the top and bottom bars of the longitudinal beams to pass through the full joint, and, therefore, these bars achieved their required development length.Cast-in-place topping slabs in the test structure were 30 mm (1.18 in.) thick and formed the diaphragms in January-February 2005 Fig. 3. Plan and elevation of test structure: (a) Plan; (b) Lateral frame; (c) Transverse frame. Dimensions in mm. Note: 1 mm - 0.0394 in. the structural system. Welded wire reinforcement (WWR) was used as reinforcement for the topping slabs. The amount of WWR ill the topping slabs was controlled by the temperature and shrinkage provisions of the MCBC. which are similar to those of AC 318-02.It is of interest to mention that the requirements for shear strength in the diaphragms given by these provisions, which are similar to those of ACI 318-89, did not control the design. A wire size of 6 x 6 in. 10/10 led to a reinforcing ratio of 0.002 in the topping slab. The strength of the WWR at yield and fracture obtained from tests were 400 and 720 MPa(58 and 104 ksi),respectively.TEST PROGRAM AND INSTRUMENTATIONTest ProgramThe test structure was subjected to simulate seismic loading in the longitudinal direction (see Fig. 3a). Quasitatic cyclic lateral loads FI and F2 were applied at the first and second levels of the structure, respectively (see Fig. 3b). The ratio of F2 to FI was held constant throughout the test., with a value equal to 2.0. This ratio represents an inverted triangular distribution of loads, which is consistent with the assumptions of most seismic codes including the MCBC.The test setup is shown in Fig. 9.The structure had Hinges A, B, and Cat each slab level as shown in Fig. 9b.The purpose of the hinges was to avoid unrealistic restrictions in the structure by allowing the ends of the slabs to rotate freely during lateral load testing. As can be seen in Fig 9, the lateral loads were applied by hydraulic actuators that work in either tension or compression.When the actuators worked in compression, they applied the loads directly at one side of the structure. However, when the actuators applied tensile loads at one side of the test structure, they were convened to compression loads at the other side by means of four high strength reinforcing bars for each actuator see D32 (1 in) reinforcing bars in Fig. 9. Both ends of these bars were attached to 50 mm (2 in.) thick steel plates At each of the floor levels, two of these plates were part of Hinge A, and the other two plates at the actuator side were part of Hinge B (see Fig. 9b).As can be seen in Fig 9b before the application of tensile loads in the actuators, the latter end of the plates left a clear space with the end of the slab. This space at zero lateral load was about 50 mm (2 in.), and it allowed for beam elongation of the structure which occurs during the formation of plastic hinges in the beams? For the case of compressive loads on the actuators acting on the transverse beam at the side of Hinge B (see Fig. 9b), the system also allowed 50 mm (2 in.) of beam elongation. These particular features of the test setup allowed the application of compressive loads at points in the slabs with no special reinforcement at these points. If tensile forces had been applied to loading points in the slabs, it is very likely that these loading points would have required unrealistic, special reinforcing details that are not represent five of those in an actual structure. The gravity load was represented by 53 steel ingots, acting at each level of the structure, with the layout shown in Fig. 9a, The weight per unit area of the ingots per level was 2.79 KPa (58.3 psf), which added to the self-weight of tile slab, leading to a total floor dead load of 5.37 KPa (112 psf). This is 88 percent of the code required gravity load for seismic design (MCBC, 1993). It was not possible to apply the remaining 12 percent of gravity load due to space limitations in the slabs. The total weight of the structure, omitting the weight of the foundation, was 284.2 KN (63.9 kips). Transverse displacements in the structure (perpendicular to the loading direction) were precluded by using steel ball hearings installed in the lateral faces of beam to-column joints of the second level at Column Lines A1and A3 (see Fig. 9a). These ball bearings were supported by a steel frame. As shown in Fig. 10, the precast columns and wails were fixed to the strong floor using steel beams supported by the foundation and anchored to the floor. The lateral loading history used in the test structure was based on force control during elastic response of the structure, followed by displacement control during yield fluctuations, using the lateral roof displacement of the structure A.The target lateral load or top displacement was typically reached by incremental loading of both actuators in which the bottom actuator was fore-contrived to half the load value of the top actuator. A cycle of lateral load of about 0.75Vg was initially applied, where VR is the nominal theoretical base shear strength computed using the ACI 318 02 provisions. The value of VR, 198 KN (44.5 kips), can be assumed to correspond to first yield in the structure.This parameter was computed using measured material properties, a strength reduction factor (1) equal to unity and common assumptions for flexural strength of reinforced concrete sections. For the lateral force of 0.75VR, the corresponding lateral roof displacement was defined as 0.75y. Using this value and assuming elastic behavior, the calculated value of the displacement at first yielding, Ny, was equal to 4.5 mm (0.18 in.). After the cycle at 0.75y, the structure was subjected to three cycles for each of the maximum prescribed roof displacements 2y, 4y, 6y, 13y and 20y-which correspond to roof drift ratios, Dr, of 0.003, 0.006, 0.009, 0.020 and 0.031, respectively. The roof drift ratio can be defined as: Dr=A/H (1) where H is the height of the structure measured from the fixed ends of first story columns. This value is equal to 2920 nun (115 in.).The complete cyclic lateral loading history applied to the structure is shown in Fig. 11. This loading history is expressed in terms of both displacement ductility, A/Ny, and roof drift ratio, Dr. In addition, the peaks of lateral displacements in Fig. 11 are related to the measured base shear force, V, expressed as the ratio V/VR.InstrumentationA detailed description of the instrumentation of the structure can be found in the report by Rodriguez and Blandon? Lateral displacements of the structure were measured with linear potentiometers placed at each level of the structure as shown in Fig. 12. Beam elongation, discussed later, was evaluated from measurements using this instrumentation. Twelve pairs of potentiometers were used for measuring the average curvatures in critical sections of two columns of the structure. In addition, nine pairs of potentiometers were instrumented in vertical sections of beams in central and lateral frames, and in sections of a wall base. Electrical resistance strain gauges were placed in the longitudinal reinforcement of some beams, columns and wails of the test structure, as well as in the hoops of beam to-column joints with substandard reinforcing details. Some of the measurements from this instrumentation and from the potentiometers in the beams are discussed in the companion paper. Here, the observed experimental response of the beam to-column joints with substandard reinforcing details is evaluated.EXPERIMENTAL RESULTSThe applied lateral loading history in the test structure is shown again in Fig.13. in this case, the peaks of lateral displacement are related to the most important events observed during testing, such as first yielding of the longitudinal reinforcement, first cracking in walls and topping slabs, loss of concrete cover, and buckling of longitudinal reinforcement.Fig. 14 shows the measured base shear force, V, versus roof drift ratio hysteretic loops. The ordinate of this graph represents the measured base shear expressed in a dimensionless form as the ratio V/VR. As shown in Fig. 14a, first yielding of the longitudinal reinforcement occurred at wails in the critical sections at foundation faces, at a roof drift ratio of about 0.0030, corresponding t9 a base shear force equal to about 1.4VR. The maximum measured base shear was equal to 549KN(123.4 kips) or 2.77VR, corresponding to a roof drift ratio of 0.020.Fig. 14b shows envelopes of interstory drift ratio, dr, measured during testing in the two levels of the structure. The results show that the two levels had similar values of interstory drifts, which is due to the significant contribution of the structural walls to the global response of the structure. This feature of the displacement profile of the structure suggests that for a given Dr Value, interstory drift values would be similar in magnitude.Cyclic lateral loading was terminated at a roof drift ratio of 0.034 corresponding to a base shear force of 462KN (103.8 kips) or 2.3Vn. At this level of lateral displacement, the buckling of longitudinal reinforcement at the fixed ends of the first-story walls was excessive, and it led to significant out-of-plane displacements of the walls along with wide cracks in the topping slab-to-wall joints as discussed in the companion paper? Fig. 14a also shows some of the most important events observed during testing. A summary of relevant damage observed during testing of die structure is described below. Wall damage was initiated by the loss of concrete cover and buckling of longitudinal reinforcement at the fixed ends of the first-story walls. These events occurred at a D, value equal to about 0.020. Buckling of the longitudinal reinforcement occurred immediately after the loss of concrete cover in these critical sections. Figs. 15 and 16 illustrate the cracking pattern and damage observed at the end of testing for the lateral frame and for the central frame, respectively. Fig. 17 shows an overall view of the damage of the lateral frame at file end of testing, and Fig. 18 provides a closer look at the buckling of longitudinal reinforcement at the fixed end of a first-story wall at the end of testing. Figs. 15 through 18 indicate that the observed damage at the end of testing in the columns and beams to-column, and beam column joints was significantly less than that of the walls.The formation of plastic hinges in the critical sections of the structural elements, such as at the fixed ends of the first story columns and wails, and in beam ends at wall faces, was observed during testing, especially after reaching the maximum base shear. Evidence of buckling of the longitudinal reinforcement was also observed in some of these sections of columns and beams (see Figs. 15 and 16), foremen of some beams, columns and wails of the test structure, as well as in the hoops of beam to-column joints with substandard reinforcing details. Some of the measurements from this instrumentation and from the potentiometers in the beams are discussed in the companion paper. Here, the observed experimental response of the beam to-column joints with substandard reinforcing details is evaluated. 一个未完工的二层预制混凝土结构物的抗震测试这篇文章是关于地震和预制混凝土建筑物设计的试验性的研究。墨西哥市里一个带有双重系统和代表了一个停车场结构的未完工的两层的预制混凝土建筑物被调查研究。这个结构物在实验室里用模拟地震荷载测试，结果失败了。在一些梁和柱的接头处，粱底部的纵筋由于尺寸的限制不能屈服。这项研究所强调的是提高所测试结构物的可观察的综合性能。这种性能表现为所测试结构物的墙控制传力途径而且能显著地减少预制结构所要求的侧向变形。源自于这项研究的预制混凝土结构抗震设计标准和规范细节被讨论。这项研究的最终结果是能更好地理解这种类型的建筑物的已得知的性能。在墨西哥，一个两层的预制混凝土构件建成的预制混凝土建筑物，在其上加上模拟的地震荷载。在这里描述的是其结果。在测试结构物中所选择的结构系统是所谓的双重类型，其定义就是构造墙的结合点以及梁-柱框架。测试结构物中预制梁柱之间的结合是窗型的。这种类型的建设显著地用在低的或中等高建筑物中，在这种建筑中在每一楼层中柱子和窗子连在一起。这些“窗”包含顶部和底部的钢筋。图1所示的是在墨西哥市中这种类型的一个商业建筑物。大多数的预制混凝土结构如图1中所示，纵梁底部的钢筋不能完全屈服。这是由于在梁-柱接头中柱的尺寸限制所造成的。为了尽力克服这种缺陷，正如在后面所描述的，在墨西哥一些工程师尝试着这样设计这些接头，就是通过用箍筋圈住这些钢筋，这样做是为了达到所要求的连续性。然而，这种尝试在ACI建筑规范和MCBC中都没有提到。这些研究的一部分是为了阐述这个观点。这项研究的目的是为了提高在实验室里的预制混凝土结构屋的可观察的性能以及为利用谕旨构件或预制结构建议了一个可接受的期望的抗震性能以及从建设能力的观点所得出的有吸引力的特征。这篇文章中强调的所测试结构物中预制构件间的连接处的可观察的性能以及预制楼层系统的性能将会详细讲述。在过去的地震中，在建筑物中造成的可观察的构造和非构造的破坏显示了通过控制结构的侧向位移来降低由地震造成的建筑物的破坏的重要性。在这里还要提到的是，在中等程度的地震中有一些情况下非结构构件的破坏相当大，尽管构造构件只有一点破坏或根本就没有破坏。这种性能和结构物中所要求的过多的侧向位移有关。为了减少地震所造成的破坏，以上的讨论建议了在结构物中可以方便地使用能控制恻向位移的构造系统。这种类型的解决方法就是所谓的双重系统。Paulay和 priestly的关于双重系统的地震反映的研究表明墙的出现降低了框架微系统中结构构件的动力要求。同时，在一个现浇的钢筋混凝土双重系统上所做的摇摆测试显示了双重系统能达到良好的抗震性能的潜力。在这次调查研究中，双重系统应用在预制混凝土构件上。双重系统的柔性要求为了使这个工程所研究的被测试结构物的能观测到的抗震反应的以后的分析打好基础，一个简单的分析模式被用来提高双重系统中主要柔性特征要求。图2所示的是一个简单的分析作用在双重系统侧向荷载反映的结果。侧向荷载从这种方式标准化，将任一系统中最大的侧向抵抗力联合起来。比如，墙和框架导致综合系统的侧向抵抗力。假设任一微系统的总的位移量为4和2。在第二种情况下，框架系统假设为弹性，墙微系统的刚度为框架微系统的4倍。图2所示，联合系统的侧向变形兼容性由墙微系统的侧向变形量控制，在第一种情况下，假设双重系统的总侧向反应有一个塑料封套，相应的位移系数是3.3在第二种情况下，框架微系统在弹性力下，起位移系数是2.5。这些简单的例子说明，在以上分析的情况下，由于在双重系统中框架微系统与墙微系统相比弹性大的多，框架微系统柔性要求更小比墙微系统的该项要求有价值。这项分析结果在被测试结构物上所做的研究上被证实了，这个证明在这篇文章的后面会讨论。有趣的是图2所示的类型的结果，Bertero在一个摇摆测试的双重系统中也发现了。测试结构物的描述在这次调查中所用到的被测试结构物是一个两层的预制混凝土建筑，是一个位于墨西哥市高地震发生地带的有代表的低层的停车结构。原型还未完工，为了简单起见，一个停车场结构物所需的扶梯在所选的结构物中没有考虑。如果考虑的话，将占有楼层系统的大面积空间，为了进行结构物的线性或非线性分析，将需要一个非常复杂的楼层系统模型。关于所测试结构物的详细的尺寸，材料，设计步骤和建设描述到处都可以发现，下面给出了这些信息的一个总结。所测试结构物的尺寸和一些特征如图3所示，其纵向以及相反方位如图3所示，同时，外部框架包含墙被定义为侧向框架，内部框架和单个T梁被定义为中间框架。纵向的两个T梁由相反方向的L型预制梁支撑如图3所示，该结构物用预制框架和预制构造墙组成，后面构件的功能是作为主要的侧向荷载抵抗系统，图4所示的是所测试结构物建设的早期阶段。我们可以看到，在柱和墙上留下了一些窗，是为了以后的预制梁的装配。墨西哥城市建筑规范所要求的设计基础剪力为0.2Wt, Wt是模型结构物的总重，假设横载为5.15Kpa，活载为0。2Kpa，模型结构物是按弹性分析的步骤设计的，比例是按MCBC要求来的，结构物中构件总的惯性都考虑了，结构物中除了中间框架的梁（会在以后介绍）以外的所有梁都考虑了刚度补偿。这些分析的结果表明测试结构物中的构造墙将承受65%的设计侧向荷载，一个用MCBC步骤考虑的结构物的名义上的侧向抵抗显示这个抵抗力是规范规定的侧向抵抗的1。3倍。这只是使结构无承载过度的因素中的其中之一，其它的以后会讨论。测试结构物的所有构件的纵筋都是从420级钢筋开始破坏的，表一是模型结构物中不同构件的混凝土压柱强度。图6，7分别是柱，构造墙和基础的钢筋详细情况，应提到的是，测试结构物是按MCBC要求设计的适度柔性结构物。由于这些规定，测试结构物不需ACI318-02第21章所要求的有边界部件的特殊的构造墙。预制的两层柱是通过埋置在一个插座连接处与预制基础相连接的基础的配筋情况以及设计步骤和性能在相应的文章中讨论。测试结构物的梁-柱接头是现浇的，为了能安置框架梁中的纵向钢筋。梁顶部钢筋是按in-situ分布在预制梁的顶部.图8所示的是中间框架中双T接头的配筋情况.因为这些T支座和支撑他们的L型梁 在轴A,C上深度相同(见图3),在双T座底部的钢筋不能穿过整个柱深,因为其被相反方向两的底不钢筋打断了.所以 ,这些带钩的钢筋只有ACI318-02第21章所要求的55%的发展长度.为了能锚固住这些带钩钢筋,在墨西哥的一些设计师沿着钩用封闭的箍筋箍住,如图8所示,这种方法的有效 性在相应的文章中回研究.测试结构物的侧向框架中的梁-柱接头中有相反方向的梁比纵梁还要深.这使的纵梁中顶部,底部的钢筋能穿过整个接头,所以这些钢筋能达到所需要的发展长度.测试结构物中顶部的现浇的板层有30mm厚,也形成了结构系统的图表.WWR被用作顶部的板层的钢筋,顶部板层中WWR的数量由MCBC中温度和收缩要求控制,这与ACI318-02中的控制要求相